Pseudo-dynamic testing of hybrid frame with steel beams bolted to CFT columns

Pseudo-dynamic testing of hybrid frame with steel beams bolted to CFT columns

Journal of Constructional Steel Research 88 (2013) 123–133 Contents lists available at SciVerse ScienceDirect Journal of Constructional Steel Resear...

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Journal of Constructional Steel Research 88 (2013) 123–133

Contents lists available at SciVerse ScienceDirect

Journal of Constructional Steel Research

Pseudo-dynamic testing of hybrid frame with steel beams bolted to CFT columns W.H. He a, Y. Xiao b, c,⁎, Y.R. Guo a, Y.L. Fan d a

State Key Laboratory of Subtropical Building Science, South China University of Technology, Guangzhou, 510640, China China Ministry of Education Key Laboratory of Building Safety and Energy Efficiency, Hunan University, Changsha, 410082, China University of Southern California, Los Angeles, CA 90089-2531, USA d College of Civil Engineering, Central South University of Forestry Science and Technology, Changsha, China b c

a r t i c l e

i n f o

Article history: Received 23 September 2012 Accepted 7 May 2013 Available online 1 June 2013 Keywords: Bolted connections Concrete filled tube Steel beam Composite slab CFT Pseudo-dynamic test Networked test Seismic behavior

a b s t r a c t This paper presents an experimental study on seismic behavior of a composite structural frame system consisting of concrete filled steel tube columns and steel beams with bolted endplate connections. Based on current seismic provisions and previous research, a ten-story prototype building was designed. Analytical models were developed to predict the elasto-plastic behavior of the prototype frame under a series of ground motion records. A four-seventh scale sub-structure model consisting of two stories and one and half span was constructed and subjected to simulated seismic excitations using pseudo-dynamic hybrid testing method. Results from the tests indicate that the model behavior under the simulated seismic loading was consistent with the expected performances analyzed for different earthquake hazard levels. Finally, quasi-static test and pushover test were also conducted to identify the ultimate failure mode of the testing model. The study shows that the proposed system can offer appropriate strength and adequate ductility required for seismic design. © 2013 Elsevier Ltd. All rights reserved.

1. Introduction Concrete filled steel tubular (CFT) column and steel beam composite structures are very efficient in load-carrying capacity and construction ability, however the complexity of its connection details and the lack of sufficient experimental validation limit its further application. Studies thus far on moment resistant CFT column and steel beam connection details have focused on exterior diaphragm and interior diaphragm connections, stiffening ring connections, beam through column connections and welded or bolted split-tee connections [1–4]. Most of these connections need considerable in-situ welding work and their seismic behaviors still need further investigation. High-strength through bolts and endplate connections are used in steel structures, which can reduce the in-situ work. Several research studies on its application in composite structures, such as steel-reinforced concrete column-steel beam connections and RC column-steel beam connections showed its promising merits in seismic design [5–7]. Experimental and analytical investigations were also performed on cross-shaped steel beam to CFT column connections [8–10]. Wu et al. tested six full size bolted connections and found those connections had excellent stiffness, strength, ductility and energy dissipation capacity. Test results showed that the model ⁎ Corresponding author at: Hunan University, Changsha, Hunan, 410082, China. Tel.: 86 731 8882 1395. E-mail addresses: [email protected].cn, [email protected] (Y. Xiao). 0143-974X/$ – see front matter © 2013 Elsevier Ltd. All rights reserved. http://dx.doi.org/10.1016/j.jcsr.2013.05.005

structure could still behave well even the story drift ratio reached 7%. Wu et al. also established a mechanical model to derive theoretical equations for calculating the stiffness, the yielding shear strength and the ultimate shear strength of the panel zone [8,9]. Limited research was also conducted on large-scale composite CFT frames, especially on those with through bolted connections [11–13]. Tsai et al. reported an experimental research on a series of pseudodynamic tests of a full-scale 3-story and 3-bay buckling-restrained braced frame (BRBF) with CFT columns. Three different types of connections were employed in their testing model, including bolted endplate connections. During the tests the bolted moment connections survived without failure after a 0.0375 radian story drift was applied where the CFT column was uplifted from the footing and all the buckling-restrained braces had fractured. Herrera et al. [13] tested a 0.6 scale 4-story 2-bay composite CFT-MRF frame with split-tee connections, of which the tee stems were fillet welded to the beam flange and the tee flange was bolted through the column. The testing results indicated that the structural performance under the simulated seismic loading was consistent with the expected performance for three different earthquake levels corresponding to the International Building Code [14]. In this study, in order to investigate the seismic performance of steel beam-CFT column frame structures with bolted endplate connections, a 10-story CFT frame was designed and a 2-story and one and half bay frame model was constructed and tested as a sub-structure of the scaled building with a scale factor of 4/7. A series of seismic simulated tests

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were conducted to study the behavior of the composite CFT frame and assess its adequacy according to current design provisions and recommendations.

shear, the peak story drift ratio and roof drift ratio of the prototype building were 2.2% and 1.4% respectively. The base shear was distributed to the interior frames and end frames according to the lateral stiffness ratio.

2. Prototype building design and analysis

2.2. Moment–rotation relationship modeling of bolted endplate connection

2.1. Prototype building design

A simplified bolted endplate connection model was developed to determine the initial stiffness of the bolted-endplate connections. As illustrated in Fig. 2a and proposed by Shi [20], an un-stiffened extended endplate can be divided into two parts for stiffness calculation, namely Part A and Part B in the figure. The initial stiffness of Part A, Kep1, is

Based on a previously completed US–Japan cooperative research program on hybrid and composite structures, a 10-story office building was designed as CFT frame structure for seismic regions. The floor plan and elevation view of a typical planar frame are shown in Fig. 1. The CFT frame was employed to provide lateral stiffness, strength and ductility for design level seismic load. As shown in Fig. 1b, all the square steel tubes were 700 × 700 mm in exterior dimension, however, with a thickness of 14 mm, 12 mm and 10 mm, respectively for different stories, resulting in different width to thickness ratios of 50, 58 and 70. The Chinese standard Q235 grade steel with a nominal yield strength fy of 235 MPa was used for the steel tube, and the nominal compressive strength fc′ of the in-filled concrete was 39 MPa. The beams were Chinese standard Q235 H-shape steel sections. CFT columns and steel beams were designed based on the requirements of the Chinese design codes [15,16] and the philosophy of “weak beam–strong column”. The design procedure of the unstiffened, bolted endplate connection followed the design methods proposed by Li et al. [7]. The prototype building was also designed to meet the requirements of current building design guidelines IBC and FEMA 450 [14,17,18]. Load combinations and equivalent lateral force procedure were determined using ASCE 7-05 [19]. The spectral acceleration at the short period (SS) and one-second (S1) was based on the deterministic limit on the maximum considered earthquake (MCE) response spectrum and equal to 1.5 g and 0.6 g respectively. According to the specified values for steel special moment resisting frames in IBC 2006, a response modification factor R = 8, an overstrength factor Ω0 = 3, a deflection amplification factor Cd = 5.5, an occupancy importance factor I = 1.0 and a story drift limit of 2.5% were selected for the building design. The base shear for drift control per IBC 2006 was 0.114W in the transverse direction, where W was the seismic weight of the building, 85,605 kN. The fundamental period of the building was 1.1 s, and the corresponding base shear for strength design was equal to 0.085W in the transverse direction, and then it was determined to be control base shear for design. Corresponding to this strength control base

K ep1 ¼ 2ðk1 þ k2 Þ

ð1Þ

where, k1 and k2 are stiffness of two segment in Part A as illustrated in Fig. 2b, and can be expressed as,

k1 ¼

e3w αe þ w 12EI GA

k2 ¼

e3f αef þ 12EI GA

!−1 ¼

!−1

e3w αew þ Eb1 t 3ep Gb1 t ep e3f

αef þ Eb2 t 3ep Gb2 t ep

¼

!−1 ð2aÞ

!−1 :

ð2bÞ

Considering the strong restraint effect of the pre-stress, the overlapping area of A1 and A2 can be considered as a four-edge supported segment, therefore the stiffness double-counted in Eqs. (2a) and (2b) is not reduced (this overlapping effect was reduced in reference [20]). Similarly, the initial stiffness of Part B, Kep2, is 3

K ep2 ¼

ef αef þ 12EI GA

!−1

3

¼

ef

Ebeff ;ep t 3ep

þ

!−1

αef

ð3Þ

Gbeff ;ep t ep

where, E is Young's modulus of steel endplate; G is shear modulus of steel endplate; I and A are the moment of inertia and area of the endplate segment cross section, respectively; α is a shear coefficient (1.5 for rectangular section); beff,ep is effectively width of endplate, as defined in Eurocode 3 [21]; tep is thickness of endplate; and b1, b2, ef,

(a) Plane view

(b) Elevation view

E

11000

9600

dxt=700x10*

H-shape steel beam

HN700x300x13x20

12000

dxt=700x12 *

8000

(3rd~10th story)

dxt=700x14*

Square CFT column

31000

9x3000=27000

32000 C

2x6400=12800

D

9600

B

HN800x300x14x26

4000

(1st~2nd story)

A Y

9600

6x6400=38400 X

1

2

3

4

5

6

12800

9600

7

Fig. 1. Schematic plot of prototype structure (all dimensions in mm). (Note:*d and t are the outside length and thickness of the square steel tube, respectively.)

W.H. He et al. / Journal of Constructional Steel Research 88 (2013) 123–133

(a) Segments division of an endplate

(b) Segments stiffness calculation

Part B

beff,ep

ef tbf ef

ef

ew ew

Part B

ef

db

Part A

125

k2

A1 Part A A2 b1 k1

b2

ew

Fig. 2. Model for determining connection initial stiffness.

ew are as defined in Fig. 2b. Thus, the initial stiffness of the endplate, Kep, can be expressed as, K ep ¼ K ep1 þ K ep2 :

ð4Þ

Since the rotation stiffness of the CFT column is much larger than that of the connection, the initial rotation stiffness of the connection, Kθ,ini, can be derived from Kep and the initial stiffness of bolts, Kbolt, which is as defined in Eurocode 3, in series and expressed as, 1

1

K θ;ini

1 ¼ þ K bolt d2b K ep d2b

ð5Þ

where, db is the distance between the centerline of tensile and compressive flange of steel beam. Based on the initial flexural stiffness given by Eq. (5) and referred to Eurocode 3 [21], a trilinear model is established, as shown in Fig. 3, to replicate the moment–rotation relationship of the connection, using the ultimate moment Mu, given by Eq. (6), of bolted endplate connection. Mu ¼ minðM ue ; M ub Þ

ð6Þ

where, Mue is the ultimate moment of endplate failure derived per yield line mechanisms and virtual work method, and Mub is the ultimate moment of bolts failure taking the prying effect into account. More detailed

Fig. 3. Moment–rotation relationship model of bolted endplate connection.

derivation procedure of these two moment capacity could be found in reference [22]. Thereafter, the yield moment was defined as 2/3 of the ultimate moment, as proposed in Eurocode 3 [21]. As illustrated in Fig. 3, the first two segments of the moment–rotation curve can be represented by Eqs. (7) and (8), whereas the last segment is a straight horizontal line that can be defined by Eq. (9), K θ ¼ K θ;ini =μ

μ¼

8 > > > <

1

1:5M j > > > : My

ð7Þ

!2:7

2 Mj ≤ Mu 3 2 M bM j ≤M u 3 u

Mj ¼ Mu :

ð8Þ

ð9Þ

2.3. Analytical investigations Nonlinear static and dynamic time-history analyses of the prototype building were conducted to evaluate its performance using the OpenSEES program [23]. All the CFT columns and steel beams were modeled as the force-based nonlinear beam-column element with discretized fiber section model. The modified uniaxial Kent–Scott– Park concrete model (Concrete02 material) was adopted to model the tensile cracking strength and linear tension softening of both confined and unconfined concrete. A constitutive model for the steel tube suggested by Fujimoto [24] and an empirical model for the steel beam developed by Muhummud [25] that accounts for the effects of local buckling in steel tubes and steel beams, respectively, were employed in the modeling. After performing a series of preliminary analyses, it is found that local buckling occurred only at the bottom segment of the first story columns and at the beam ends in the lower several stories, while the other structural members exhibited only limited inelastic deformations well below the deformations at which local buckling would initiate. Therefore, the prototype building model included local buckling model at the members mentioned above only. The Menegotto–Pinto steel material (Steel02 material) with isotropic and kinematic strain hardening was used for the other steel beam and steel tube elements. A panel zone strength model proposed by Wu et al. with considering the contribution of the endplate stiffness was utilized to calculate the shear capacity of the panel zone [8]. In the analytical model, the shear-deformation behavior of the panel zone was represented by a 6-node model, in which included four boundary nodes and two master nodes as shown in Fig. 4. The boundary node

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Test OpenSEES

Fig. 6. Comparison of analysis model prediction with experimental response of panel zone specimen FSBE-8 in Wu et al. [8]. Fig. 4. Moment–rotation relationship model of bolted endplate connection.

displacements and rotations are appropriately slaved to those of the two master nodes M1 and M2. These two master nodes were connected with a non-length element to make their displacements identical and their rotations independent. At each end of the steel beam, a series of rotational spring and shear spring was set up to model the nonlinearity of the bolted endplate, as shown in Fig. 5. The stiffness of rotation spring Kθ could be calculated by Eq. (7). The stiffness of shear spring Kv was equal to the shear stiffness of the bolts. Cyclic analysis on connection element was exercised to verify the accuracy of the model compared with previous test results. It is evident that the hysteretic behavior of connection simulated by OpenSEES illustrated in Fig. 6 is positive and agrees well with the experimental results obtained in reference [8]. The frame model presented in this paper utilizes the measured material properties of steel beams, CFT tubes, and filled-in concrete for the CFT columns as shown in Table 1. A uniform lateral load pattern over the height of the building and a lateral load pattern based on the recommendations from IBC 2006 were used to perform the pushover analyses. Fig. 7a shows the relationships between the normalized base shear (base shear divided by seismic weight W) and the normalized roof displacement (NRD, roof displacement

divided by building height) obtained from the pushover analyses. In Fig. 7b, the performance is compared with demand calculated based on ATC-40 [26] capacity-spectrum method for the IBC 2006 lateral load profile. A spectral performance displacement of 549 mm was obtained corresponding to the MCE level nonlinear response spectrum, which was 735 mm after converted to actual roof displacement and resulted in an NRD of 2.4%. The prototype building performance under three ground motion intensity levels was investigated, including the frequent occurrence earthquake (FOE), the design bass earthquake (DBE) and the maximum credible earthquake (MCE), corresponding to the probability of exceedance of 50%, 10% and 2% in 50 years, respectively [18]. The performance objectives and related limit states were based on those defined in NEHRP recommendation provisions [17] for three performance levels, namely operational, life safety and collapse prevention. Based on previous research on classifying natural ground motions [27], a total of 15 ground motion records were selected from the Pacific Earthquake Engineering Research Center (PEER) acceleration database [28], with magnitudes ranging from 6.5 to 7.4, durations ranging from 24.4 s to 44.0 s and scale factors of MCE input level ranging from 0.878 to 3.699, respectively. The 15 ground motions were scaled so that their average value of the 5% damped response spectra is not less than the design response spectrum for periods ranging from 0.2T to 1.5T, here T is the fundamental period of the prototype building. Time-history analyses were conducted using these records scaled to three seismic input levels. Because the selected ground acceleration records represent only a sample of all the possible records that can affect the building, the performance evaluation presented was carried out using statistical values. The median and 84th-percentile values for each response quantity were included, where these values represented 50% and 16% probability of exceedance, respectively. Based on the analyses, it is confirmed that none of the story drifts corresponding to DBE and MCE earthquake input levels exceeds 2%, indicating the adequacy of the prototype building design.

Table 1 Material properties of test sub-structure.

Fig. 5. Simplified connection model.

Location

Steel σy (MPa)

σu (MPa)

εy (%)

εu (%)

Flange Web Tube

292 340 319

448 476 432

0.20 0.23 0.19

17.21 14.62 11.47

Concrete fc′ (MPa) 16.7 (slab concrete) 54.4

W.H. He et al. / Journal of Constructional Steel Research 88 (2013) 123–133

(a) Normalized Base Shear vs. NRD

127

(b) Demand Spectrum vs. Capacity Spectrum

0.7 1.6

Spectral Acceleration / g

Normalized base shear

0.6 0.5 0.4

Concrete cracking CFT yielding Steel beam yielding CFT local buckling Uniform load profile IBC 2006 load profile IBC 2006 design base shear

0.3 0.2

Elastic Spectrum for MCE Nonlinear Demand Spectrum

1.2

Capacity Spectrum

0.8

Equivalent Binlinear Capacity Spectrum SRa=0.50 SRv=0.61 Perfomance Point=549mm

0.4

0.1

0.0 0

0

0

1

2

3

4

400

800

1200

1600

Spectral Displacment / mm

5

Normalized displacement (%) Fig. 7. Pushover analysis results of prototype building.

4. Test set-up and loading program As shown in Fig. 10, the test sub-structure was fixed on the strong reaction floor with high-strength steel bolts. Lateral displacements were imposed to the model through two 1000 kN hydraulic actuators

6857

dxt=400x7 *

4571

dxt=400x8*

HN400x200x8x13 (3rd~10th story)

Test substructure (See Fig. 7)

Analytical substructure

0

2286

HN450x200x9x14 (1st~2nd story)

17714

9x1714=15426

6286

The prototype building was scaled with a factor λ of 4/7 and divided into two parts, i.e., the test sub-structure and the analytical sub-structure in the hybrid experimental study, as schematically shown in Fig. 8. The test sub-structure was a model of one and a half bay of the bottom two stories of the scaled building, including two heavily reinforced rigid footings which were designed according to ACI 318-02 [29], as shown in Fig. 9. The upper eight stories, or the analytical sub-structure, was modeled by tri-linear model, in which the stiffness of each story was obtained from pushover analyses with the IBC 2006 lateral load pattern, with lumped mass in the general networked testing platform NetSLab (Network Structural Laboratories) developed by Xiao et al. [30,31] for remotely cooperative hybrid testing. The main details of the test substructure are shown in Figs. 10 and 11. The test sub-structure was made of two columns and two floors forming one and a half span with length of 5.49 m and 3.66 m, respectively, and a total height of 4.0 m. Each structural floor was a composite beam with a cast-in-place slab, which was 87 mm thick and 1232 mm wide. The steel beams were connected with the slab by 16 mm diameter shear studs, spaced at

140 mm. The floor slabs were reinforced with double layer reinforcement mats of 10 mm deformed bars spaced at 150 mm. The steel tubes of the CFT columns were manufactured by welding two cold-form U-shape Q235 steel plates of 8 mm thickness by full penetration welding, forming a hollow square section of 400 × 400. All the beams were Chinese standard H-shape Q235 steel HN450 × 200 × 9 × 14 (section height h = 450 mm, flange width bf = 200 mm, web thickness tw = 9 mm, flange thickness tf = 14 mm). The end-plate was welded to each end of the beams with double-bevel groove welds and connected to the CFT columns with high-strength unbonded threaded bolts. The details of the connections are shown in Fig. 12. As the strength of the concrete filled in the steel tube was fully developed, a prestress of 290 kN was applied to each bolt.

dxt=400x6 *

3. Test model

5486

7314

5486

Note:*d and t are the outside length and thickness of the square steel tube, respectively. Fig. 8. Schematic plot of test structure. Note:*d and t are the outside length and thickness of the square steel tube, respectively.

128

W.H. He et al. / Journal of Constructional Steel Research 88 (2013) 123–133

Fig. 9. Footing construction.

at each story. According to the base shear force method and the configuration of the one-half span model, the load imposed by the lower actuator would be half of the story force of the first story, while the horizontal load imposed by the upper actuator would be the half of the total of the story forces above the first story. In order to distribute the larger force of the upper actuator more evenly to the second story, two steel beams were designed as loading arms and attached to the concrete slab of the upper floor slab and connected with the actuator, as shown in Fig. 11. Simple pins were designed to connect the end point of the half span floor beams, where the inflection points were assumed, to eliminate possible errors due to the different rotation angles of the two actuators. Because the axial load caused by overturning moment was only about 10% of the gravity loads for the first story columns through primary estimation, the axial load applied to each column was determined to be kept constant during testing for simplification. Constant axial loads with the axial load ratios of 0.38 and 0.23 were applied to the interior and the exterior columns, respectively, through post-tensioning 55 mm diameter high-strength steel rods using two 1500 kN capacity hydraulic hollow jacks. The forces of the rods were transferred to each column by a cross beam mounted on top of the column. To eliminate the bending of the high-strength rods, a specially designed pin was connected to the lower end of each rod. The imposed lateral displacement was measured by both the displacement transducer of the actuator and linear potentiometers. The

High strength rod Exterior column

2286

corresponding lateral forces were recorded by the built-in load cells of the actuators. Inclinometers measured rotations across the potential plastic hinge regions in the beams and the first story CFT columns, as well as rotations along the bolted beam-to-column connections. Electrical resistance strain gauges were affixed on the surface of the steel tube near the base of the column and on the beams and columns near the connections.

N

S

Reaction wall

Cross beam

Slab thickness=86

1714

200

2x1500kN Hollow jackets

Fig. 11. Photo of test set-up.

H-shape steel beam HN450x200x9x14 CFT column 400x400x8 Footing 2000x1500x400

5486

Pin connections Interior column

2x1000kN Actuators

Bolted connection Pin Reaction floor

3657 Fig. 10. Schematic of test set-up.

(All dimensions in mm)

W.H. He et al. / Journal of Constructional Steel Research 88 (2013) 123–133

quasi-static loading test, the displacements imposed to the first story of the test structure followed the loading protocol suggested by AISC 341-05 [31] while the displacements of the second story were calculated from the first mode based on the displacements of the first story and the experimentally determined initial stiffness matrix of the test substructure. In order to investigate the maximum responses of the structural members under the available loading capacity of the actuators, the pushover test was performed till the story drift ratio of the first story exceeded 6%, even after the loading capacity of the second story actuator was almost fully utilized. It is to be noted that though in both the quasi-static test and the pushover test, the shear force proportion of the first story and the second story, as well as the gravity loads, was not fully coordinate as that of the prototype building, the responses of these two tests still had reference value for performance evaluation of the suggested structure system.

54 MPa concrete

25

Q235 steel beam 450x200x9x14

CFT column 400x400x8

Q235 steel tube

129

Unbonded 10.9G threaded M27 bolt

5. Experimental response Fig. 12. Typical connection in scaled frame.

Feed-back displacements of the test sub-structure in the four PSD tests are exhibited in Fig. 14. Envelopes of story drift and shear force at each floor are presented in Fig. 15(a) and (b), respectively. The main test sub-structure responses of each test are summarized in Table 4. In the calculation of the overstrength factor, the design base shear was based on that of a single interior frame of the prototype building. Base shear distributed to this frame is calculated according to the ratio of the stiffness of the single frame to that of the whole structure in transverse direction, and then is scaled by λ 2 and equaled to 330 kN, where λ is the scale factor of the test structure and equaled to 4/7. Results from each test are discussed in the following sessions.

Table 2 Input virtual lumped mass. Story

1

2

3

4

5

6

7

8

9

10

Mass (kN)

504

389

377

377

377

376

376

376

376

354

Earthquake simulation tests were performed utilizing quasi-static test procedure and sub-structure pseudo-dynamic test procedure which combines numerical computation, online control and experimental measurement. The seismic mass matrix M and damping matrix C were analytically defined. The seismic mass M for the scaled structure was based on the analysis results of the prototype building listed in Table 2. The damping matrix C was a Rayleigh damping matrix, based on a damping ratio of 5%, the analytical mass matrix M and a hybrid stiffness matrix K. The hybrid stiffness matrix was composed of test sub-structure stiffness matrix KT which was twice of the experimentally determined initial elastic stiffness matrix, and analytical sub-structure stiffness matrix KA which was derived based on the trilinear model for each story from the pushover analysis results of the scaled structure. Due to only limited inelastic deformations exhibited in upper eight stories from the primary analyses, the residual non-linear responses developed in those stories were neglected. The NetSLab platform adopted the α method [30], an unconditionally stable implicit time integration method, to solve the governing equations of motion with a time step of 0.02 s. A small tolerance of 0.1 mm was used to ensure the imposed displacements being sufficiently close to the target displacements up to drift level of 0.5%. Thereafter, the tolerance was increased to 0.5 mm. A total of six loading excursions were executed, including four pseudo-dynamic (PSD) tests, a quasi-static test and a pushover test, as shown in Table 3. According to the time-history analysis results, the ground motion record IVEL shown in Fig. 13(a) was selected and scaled to represent the seismic input for FOE and DBE events, and KBNA shown in Fig. 13(b) was selected and scaled to represent the seismic input for MCE and MCE-after events, with respective scale factors. During the

5.1. FOE loading level The maximum roof displacement during the FOE simulation was 30.1 mm, corresponding to an NRD of 0.17%. The maximum story drift ratio was 0.23%, occurring in the 4th story. The response of the frame was nearly linear elastic. The maximum base shear of the test structure was 417 kN, or 11% larger than the design base shear. The energy dissipated was 4 kN·m. During the FOE test, only a small amount of noticeable cracks in the first story concrete slab were observed, indicating that the target performance level was reached by the test structure. 5.2. DBE loading level The peak story drift of the DBE event reached 1.11% in the second story, and the maximum roof displacement was 105.6 mm, or an NRD of 0.6%. Only minor inelastic deformations were induced in the test structure. The maximum base shear attained by the test structure was 1598 kN, or 4.27 times the design base shear, as seen in Table 4. Cracks in the concrete slab of the first story near the interior connection developed, and the measured strains of the steel tubes and steel beams near their ends were detected exceeding the yield strain. However, no obvious local buckling and strength degradation were developed in the model structure.

Table 3 Test matrix. Test name

Description

Test method

Scale factor (SF)

PGA

Record

FOE DBE MCE MCE after Cyclic Pushover

FOE level DBE level MCE level MCE after level AISC loading protocol Pushover

Pseudo-dynamic Pseudo-dynamic Pseudo-dynamic Pseudo-dynamic Quasi-static Quasi-static

SF SF SF SF – –

0.14 g 0.48 g 1.01 g 1.63 g Maximum roof drift ratio = 1.6% Maximum story drift ratio = 6%

IVEL IVEL KBNA KBNA AISC 341-05 –

= = = =

0.447 1.523 1.976 3.188

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W.H. He et al. / Journal of Constructional Steel Research 88 (2013) 123–133

(a) IVEL Acceleration (g)

0.6 0.4 0.2 0 -0.2 0

5

10

15

20

25

30

20

25

30

-0.4 -0.6

t (s)

(b) KBNA Acceleration (g)

0.6 0.4 0.2 0 -0.2 0

5

10

15

-0.4 -0.6

t (s) Fig. 13. Selected ground motion records for PSD tests.

expected for the model structure subjecting to the MCE loading was reached.

5.3. MCE loading level The maximum roof displacement in MCE test was 120.4 mm corresponding to an NRD of 0.68%. The peak story drift ratio ranged from 0.46% to 1.17% in the structure and the largest value occurred in the 2nd story. The maximum base shear reached 1809 kN with the overstrength factor Ω0 being 4.84. Extensive yielding was developed in both the steel beam and the CFT column end regions. Local buckling was found at the bottom end of the exterior column in the first story. More cracks in the concrete slab were observed with a maximum width of approximately 0.4 mm. The stiffness of the test sub-structure was degraded compared with the initial stiffness measured before the MCE test. However, the stiffness degradation did not severely impair the global stability of the structure. It is considered that the limit state

5.4. MCE-after loading level The loading case MCE-after is a further excursion beyond the MCE loading level. The recorded maximum displacement at the roof was 185.6 mm or an NED of 1.05%. The maximum drift θmax was 1.80% occurring in the second story. As presented in Fig. 16, the test structure showed limited hysteretic response with a maximum base shear of 2176 kN, corresponding to an overstrength factor Ω0 of 5.82. The energy dissipated was 353 kN · m, which was six times of the MCE test. Inelastic deformations developed in the beams where yielding was concentrated at the beam ends over a length of approximately two-thirds of

50 1st story

40

2nd story

Lateral Displacement / mm

30 20 10 0 -10 -20 -30 -40 -50

0

FOE

t=0~30s

30

DBE

t=0~30s

60

MCE

t=0~30s

90

t/s Fig. 14. Displacement responses of test sub-structure in pseudo-dynamic tests.

MCE-after t=0~30s

120

W.H. He et al. / Journal of Constructional Steel Research 88 (2013) 123–133

(a) Story drift ratio

131

3000

10 MCE-after

2000

9

Quasi-static

Base shear / kN

8 7

Story

6 5 FOE

4

DBE

3

MCE MCE-After

1000 0 -1000 -2000

2 1

-3000

0 0

0.5

1

1.5

2

-2

-1.5

-1

-0.5

0

0.5

1

1.5

2

Story drfit ratio/ %

Story drift ratio (%) Fig. 16. First story hysteretic response in MCE-after and quasi-static tests.

(b) Story shear was calculated with a proportion of 1.76 to the first story based on the fundamental mode analysis. The peak story drift ratio reached 1.64% in the second story. The maximum base shear was 2533 kN. The base shear-story drift ratio hysteretic response of the first story presented in Fig. 16 indicates that the structure had not reached its full capacity.

10 9 8 7

Story

6

5.6. Pushover test

5 FOE

4

DBE

3

MCE MCE-After

2 1 0 0

1000

2000

3000

Story shear (kN) Fig. 15. Envelop of global response in PSD tests.

the beam height from the end-plate. Local buckling of the steel tube developed in both the exterior and interior columns. Throughout the MCE-after test, concrete floor slabs were found crushed around the beam-to-column connection, but structural damage of the CFT frame was not significantly severe. 5.5. Quasi-static test The imposed displacement of the first story followed the loading protocol proposed by AISC 341-05, while the loading history of the 2nd story Table 4 Response of test sub-structure.a Test

θmax (%)

θmax,h (%)

θmax,col (%)

θmax,bm (%)

θmax,conn (%)

Ed (kN·m)

Ω0

FOE DBE MCE MCE after Quasi-static Pushover

0.23 1.11 1.17 1.80 1.64 5.96

0.17 0.60 0.68 1.05 1.63 4.56

0.13 0.22 0.43 5.27 5.01 6.25

0.11 0.22 0.28 2.16 2.31 3.62

0.11 0.18 0.21 1.03 0.78 1.82

4 43 58 353 402 428

1.11 4.27 4.84 5.82 6.78b 9.68b

a θmax, θmax,h, θmax,bm, and θmax,conn, are the maximum story drift ratio among all stories, maximum roof drift ratio, maximum beam total rotation, maximum column total rotation, maximum connection rotation, Ed, total energy dissipation of the test sub-structure, and Ω0 is overstrength factor of the test structure, respectively. b These two values listed here were for reference only since loading pattern was different from the PSD tests.

In order to fully examine the capacity of the model structure, pushover loading test was finally executed, which consisted of two stages. During the first loading stage, displacements imposed to both stories followed the same loading ratio as the quasi-static test. After the hydraulic actuator for the second story loading reached 93% of its capacity, the lateral force provided by this actuator was then held constantly for safety consideration, whereas only the first story actuator was continuously loaded. The base shear-story drift ratio response of the first story is illustrated in Fig. 17, which indicates that the test structure had excellent ductility. After the story drift ratio of the first story reached 6%, severe local buckling in the web near the exterior end of the steel beam was observed, along with the endplate being pried away slightly from the steel tube in the first story as shown in Fig. 18. Out-of-plane local buckling of the square tube developed obviously in its compression face near column end at the 5% first story drift ratio, as shown in Fig. 19. The test was terminated when the peak drift ratio 6% of the first story was reached. The maximum base shear in the pushover test reached 3621 kN. The maximum rotation of the beams, columns and connections was significantly larger than those in the PSD tests, as shown in Table 4. 6. Comparison between tests and analyses The comparison of maximum story shear between PSD tests and analyses using OpenSEES is exhibited in Fig. 20. For most cases, the analytical story shears for the first and second stories are slightly larger than those obtained from the tests. Overall, the analytical results are reasonably close to the test results, indicating that the modeling method described in this paper predicated the behaviors of the tests reasonably well. 7. Discussion on design criteria and performance As shown in Fig. 15a, in the four PSD tests, none of the story drifts exceeds the IBC 2006 design drift limit of 2.0%. The test observations also showed no significant strength degradation. The story shear envelopes of four PSD tests shown in Fig. 15b correspond to over-strength factors of 1.11, 4.27, 4.84, and 5.82 for the FOE, DBE, MCE, and MCE-after tests,

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4000 3500

Base shear / kN

3000 2500 2000 1500 1000 Pushover

500 0

0

1

2

3

4

5

6

Story drfit ratio/ % Fig. 17. First story response in pushover test.

Fig. 19. Local buckling at column end.

respectively. The peak story shear of the FOE slightly exceeded the design story shears, while during the other tests, it was significantly larger than the design values. For the DBE and MCE tests, the overstrength factors are 1.42 and 1.63 times the IBC 2006 recommended value of 3.0 for special MRFs, respectively, indicating that an increase in overstrength factor may be needed to ensure adequate structural safety under severe earthquake. As presented in Table 4 and observed during the tests, the deformations were concentrated in the ends and at the bottom ends of the CFT columns in the first story. Strain gauge readings and experimental observations show that yielding and deformation in the beams were more significant than those of columns. Therefore the test structure realized the weak beam–strong column design intention. The maximum rotation of the beam-to-column connection occurred in the pushover test was only 1.82% corresponding to the story drift ratio of 6%, which met the design criteria proposed by AISC 341-05. And hence, the structure also complied with the design philosophy of weak member–strong connection. During the pushover test, although the imposed displacement exceeded the IBC 2006 target drift ratio of 2.0% and even reached 6%, the beams, columns and beam-to-column connections still had considerable ductility.

prototype building. Four pseudo-dynamic tests were conducted to simulate different levels of seismic inputs. The performance objectives for the structure included operational, life safety and collapse prevention performance level for the FOE, DBE and MCE, respectively. Low cycle reversed quasi-static loading test and pushover test were also conducted to study the possible ultimate behavior of the structure. The test structure responses indicated that the composite steel beam-CFT column structure can achieve excellent seismic performance. The expected performance level was reached in the corresponding pseudo-dynamic tests. Inelastic deformations were concentrated at the ends of the beams and the bottom ends of the first story columns. The bolted endplate connections with high-strength steel bolts performed extremely well. The design philosophy of weak beam–strong column and weak member–strong connection was realized. It is found that the overstrength factor Ω0 = 3.0, provided by current seismic provisions for special moment resisting frames, was relatively small for the structure studied herein and larger values should be considered in future design. The experimental results were reasonably well simulated with the analytical models developed for the CFT frames with composite steel beam with bolted connections. The pseudo-dynamic tests conducted in this work also verified the usefulness and effectiveness of the networked structural testing platform NetSlab.

8. Conclusions

Acknowledgments

In order to investigate the seismic behavior of composite frame with steel beams bolted to CFT columns, a 4/7 scale sub-structure model was designed and tested per the performance based design of a 10-story

The research project was carried out at the Ministry of Education Key Laboratory of Building Safety and Energy Efficiency and the Center for Integrated Protection Research of Engineering Structures (CIPRES)

Fig. 18. Web distortion and end-plate deformation.

Fig. 20. Comparison of test and analysis.

W.H. He et al. / Journal of Constructional Steel Research 88 (2013) 123–133

of the Hunan University. The research was sponsored by the National Natural Science Foundation of China (NSFC no. 90715036, and 51161120360) and the Program for Changjiang Scholars and Innovative Research Team Project from the Ministry of Education of China (no. IRT0619). The international collaboration was facilitated by the China Scholarship Council (CSC) and the University of Southern California. The authors would particularly like to thank the anonymous reviewer whose very knowledgeable comments have significantly helped the authors in finalizing the paper. References [1] Alostaz YM, Schneider SP. Analytical behavior of connections to concrete-filled steels tubes. J Constr Steel Res 1996;40(2):95–127. [2] Schneider SP, Alostaz YM. Experimental behavior of connections to concrete-filled steel tubes. J Constr Steel Res 1998;45(3):321–52. [3] Elremaily A, Azizinamini A. Design provisions for connections between steel beams and concrete filled tube columns. J Constr Steel Res 2001;57:971–95. [4] Ricles JM, Peng SW, Lu LW. Seismic behavior of composite concrete filled steel tube column wide flange beam moment connections. J Struct Eng 2004;130(4): 223–32. [5] Chou CC, Uang CM. Cyclic performance of a type of steel beam to steel-encased reinforced concrete column moment connection. J Constr Steel Res 2002;58(5–8): 637–63. [6] Li X, Wu YT, Mao WF, Xiao Y, Anderson JC, Guo YR. Bolted end plate connections for steel reinforced concrete composite structures. Struct Eng Mech 2006;24(3): 291–306. [7] Li X, Xiao Y, Wu YT. Seismic behavior of exterior connections with steel beams bolted to CFT columns [J]. J Const Steel Res 2009;65:1438–46. [8] Wu LY, Chung LL, Tsai SF, Lu CF, Huang GL. Seismic behavior of bidirectional bolted connections for CFT columns and H-beams. Eng Struct 2007;29(3):395–407. [9] Wu LY, Chung LL, Tsai SF, Shen TJ, Huang GL. Seismic behavior of bolted beam-to-column connections for concrete filled steel tube. J Const Steel Res 2005;61(10):1387–410. [10] Ricles JM, Peng SW, Lu LW. Seismic behavior of composite concrete filled steel tube column-wide flange beam moment connections. J Struct Eng ASCE 2004;130(2): 223–32. [11] Tsai KC, Hsiao PC, Wang KJ, et al. Pseudo-dynamic tests of a full-scale CFT/BRB frame — part I: specimen design, experiment and analysis. Earthq Eng Struct Dyn 2008;37(7):1081–98. [12] Tsai KC, Hsiao PC. Pseudo-dynamic test of a full-scale CFT/BRB frame — part II: seismic performance of buckling-restrained braces and connections. Earthq Eng Struct Dyn 2008;37(7):1099–115. [13] Herrera RA, Ricles JM, Sause R. Seismic performance evaluation of a large-scale composite MRF using pseudo-dynamic testing. J Struct Eng ASCE 2008;134(2): 279–88.

133

[14] International Code Council (ICC). International Building Code 2006. Falls Church: ICC Inc.; 2006. [15] China Association for Engineering Construction Standardization. Technical specification for structure with concrete-filled rectangular steel tube members.CECS159. Beijing: China Planning Press; 2004 [ISBN: 1580058588 In Chinese]. [16] China Ministry of Construction. Code for design of steel structures.GB 50017. Beijing: China Planning Press; 2003 [ISBN: 1580058536 In Chinese]. [17] Federal Emergency Management Agency (FEMA). NEHRP recommended provisions for new buildings and other structures. Part 1—provisions. Washington, D.C.: Rep. no. FEMA 450; 2003 [18] Federal Emergency Management Agency (FEMA). NEHRP recommended provisions for new buildings and other structures. Part 2—commentary. Washington, D.C.: Rep. no. FEMA 450; 2003 [19] American Society of Civil Engineers (ASCE). Minimum design loads for buildings and other structures. ASCE/SEI 7-05. Reston, Va; 2005. [20] Shi G. Static and seismic behavior of semirigid end-plate connections in steel frames. Ph.D. dissertation Beijing, China: Department of Civil Engineering, Qinghua University; 2004 [In Chinese]. [21] European Committee for Standardisation (CEN). Eurocode 3. Design of steel structures, part 1–8: design of joints (EN 1993-1-8:2005). Brussels; 2005. [22] Mofid M, Ghorbani AM, McCabe SL. On the analytical model of beam-to-column semi-rigid connections, using plate theory. J Thin-walled Struct 2001;39:307–25. [23] Mazzoni S, McKenna F, Fenves GL. Open system for earthquake engineering simulation user command-language manual [OL]. California: Berkeley http://opensees. berkeley.edu/OpenSees/manuals/usermanual/index.html . [Last visit in June 2008]. [24] Fujimoto Toshiaki, Mukai Akiyoshi, Nishiyama Isao, et al. Behavior of eccentrically loaded concrete-filled steel tubular columns. ASCE J Struct Eng 2004;130(2): 203–12. [25] Muhummud T. Seismic design and behavior of composite moment resisting frame constructed of CFT columns and WF beams [D]. Bethlehem: Lehigh University; 2004110–35. [26] Applied Technology Council (ATC). Seismic evaluation and retrofit of concrete buildings (ATC-40). Report no. SSC 96-01. Redwood City: C.A., 1996. [27] Seo CY. Influence of ground motion characteristics and structural parameters on seismic behavior of SDOF system. Ph.D. dissertation Bethlehem, PA: Civil and Environmental Engineering Department, Lehigh University; 2004. [28] Pacific Earthquake Engineering Research Center (PEER). Strong motion database.Berkeley: PEER; 2007 [http://peer.berkeley.edu/smcat/index.html, Last Visit in June 2007]. [29] American Concrete Institute (ACI). Building code requirements for structural concrete (ACI 318-02) and commentary (ACI 318R-02). Farmington Hills: Mich, 2002. [30] Xiao Y, Hu Q, Guo YR, Zhu PS, Yi WJ. Network platform for remote structural testing and shared use of laboratories. Prog Nat Sci 2005;15(1):1135–42. [31] Xiao Y, Guo YR, Zhu PS, Kunnath S, Martin GR. Networked pseudodynamic testing of bridge pier and precast pile foundation. Eng Struct May 2012;38:32–41.